Evaluation on stability condition along the headrace tunnel of Kulekhani-III Hydroelectric Project

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Nitesh ShresthaEvaluation on stability condition along the headrace tunnel of Kulekhani-III Hydroelectric Project NTNU Norwegian University of Science and Technology Faculty of Engineering Department of Geoscience and Petroleum

Master ’s thesis

Evaluation on stability condition along the headrace tunnel of Kulekhani-III Hydroelectric Project

Master’s thesis in Hydropower Development Supervisor: Krishna Kanta Panthi

June 2020

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Evaluation on stability condition along the headrace tunnel of Kulekhani-III Hydroelectric Project

Master’s thesis in Hydropower Development Supervisor: Krishna Kanta Panthi

June 2020

Norwegian University of Science and Technology Faculty of Engineering

Department of Geoscience and Petroleum

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NTNU Department of Geoscience and Petroleum

Norwegian University of Science and Technology

Your ref.: MS/I19T55/IGP/NSKP Date: 06.01.2020

TGB4910 Rock Engineering - MSc thesis for

Nitesh Shrestha

EVALUATION ON STABILITY CONDITION ALONG THE HEADRACE TUNNEL OF KULEKHANI III HYDROELECTRIC PROJECT

Background

Himalayan rock mass are subjected to extensive earthquakes caused by tectonic movement. This causes extensive fracturing in the rock mass near the surface and at the rock mass of major tectonic faults and adjacent areas. The headrace tunnel of the Kulekhani III hydroelectric project is located between the Main Boundary Thrust (MBT) and Main Central Thrust (MCT) of the Himalaya. The rock mass at the area are highly fractured and deformed which caused considerable amount of block falls and also plastic deformation during tunnel excavation. The candidate was involved at the project during his project work and has access to data and information about the project.

MSc thesis task

Hence, this MSc thesis is to focus on the documentation and evaluation of block falls and plastic deformation along the headrace tunnel of the Kulekhani III Hydroelectric Project, with a focus on the following issues:

 Review existing theory on the stability issues in underground excavation with focus on the block falls and plastic deformation.

 Briefly describe about Kulekhani III Hydroelectric Project. Discuss about the extent of engineering geological investigations carried out at the project.

 Assess predicted and actual rock mass conditions as well as rock support use along the tunnel alignment.

 Evaluate block falls and plastic deformation using prevailing rock engineering theory. While carrying out evaluation on block falls consider seismic load.

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 Analyse block falls using wedge unwedge program considering seismic load. Analyse plastic deformation using numerical modelling for the selected segments of the headrace tunnel.

 Compare and discuss the analysis results from analytical and numerical approaches.

Relevant computer software packages

Candidate shall use rocscience package and other relevant computer software for the master study.

Background information for the study

 Relevant information about the project such as reports, maps, information and data collected by the candidate.

 Scientific papers, reports and books related to the Himalayan geology and tunnelling.

 Scientific papers and books related to international tunnelling cases.

 Literatures in rock engineering, rock support principles, rock mechanics and tunnelling.

Mr. Bibek Neupane will be the co-supervisor of this MSc thesis.

The thesis work is to start on January 15, 2020 and to be completed by June 10, 2020.

The Norwegian University of Science and Technology (NTNU) Department of Geoscience and Petroleum (IGP)

January 06, 2020

Dr. Krishna Kanta Panthi

Professor of rock and tunnel engineering, main supervisor

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This Master thesis entitled“Evaluation on stability condition along the headrace tunnel of Kulekhani-III hydroelectric project”is submitted to Department of Geoscience and Petroleum at Norwegian University of Science and Technology (NTNU) in partial fulfillment of the requirement for Master of Science in Hydropower Development (2018- 2020) under Department of Civil and Environmental Engineering, NTNU.

The thesis is focused on evaluation of wedge/block failures and plastic deformation experienced at Kulekhani-III hydroelectric project located at lesser Himalayan region of Nepal. Analysis is done with different methods such as empirical, semi-empirical, analytical and numerical methods based on data collected from Kulekhani-III Hydroelectric project, Nepal. The thesis work started during spring semester of 2020 and is submitted at the end of spring semester of 2020.

- - - - Nitesh Shrestha NTNU, Trondheim June, 2020

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I would like to express my earnest gratitude and deep respect to my supervisor, Profes- sor Dr. Krishna Kanta Panthi for his constant availability, encouragements, motivations regarding the project, valuable suggestions and sharing of ideas and pushing me to my potential and making this project what it is now without which this thesis would not have been completed. I thank PhD fellow Bibek Neupane, co-supervisor to me who contributed to brainstorming and discussions for guiding through thesis.

I express my gratitude to Nepal Electricity Authority (NEA) who granted me to conduct my project work at Kulekhani-III Hydroelectric Project. I would like to thank Kulekhani- III Hydroelectric Project, Sanutar for supporting me and providing me all the possible documents required for carrying out this project. I am grateful to Er. Basanta Bahadur Singh and Er. Arpan Bahadur Singh, assistant project managers at Kulekhani-III Hydro- electric Project for providing me necessary supporting geological documents for carrying out this thesis. I am also thankful to consultants from Sino-Hydro for providing me the geological data of Kulekhani-III Hydroelectric Project. My special thanks goes to Dr. Niga Shrestha for her kind proofreading.

I am grateful to Professor Oddbjørn Bruland, in-charge and coordinator for master program on Hydropower development at NTNU for his support and motivation to study and complete this master degree. I am thankful to Norwegian Agency for Development Co- operation (NORAD) for the scholarship to support my living during my stay at Trondheim.

I am deeply obliged to my parents for their never ending support and encouragement.

I dedicate this work to my parents.

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With over 90% of the country’s total electricity generation capacity generated by hydropower plants, Nepal is heavily dependent on hydro resources to meet its energy de- mands. Best use of water using the shortest and the most reliable water ways through steep mountain terrains with underground tunnels can help harness this gift of nature in the most efficient way. However, there are numerous risks and uncertainties involved with underground construction works. The major technical challenges involved in underground tunneling are stress induced instabilities, water pressure and overwhelming cost during construction. In case of Himalaya, due to active tectonic movements, rock masses are highly stressed and fragile. They are not capable to withstand high in-situ stress. These rocks are soft and plastic. Most often, block/wedge failures and plastic deformation are instability issues in these rock masses.

Block/wedge failures and plastic deformation analysis have been carried out with reference to tunnel instabilities issues witnessed along headrace tunnel (HRT) at Kulekhani-III hydroelectric project (KL-III HEP). HRT is constructed along rock masses such as mar- ble, schist, quartzite, phyllite and dolomite. In addition to that, the tunnel crosses an active major thrust (Mahabharat thrust) and has been exposed to considerable amount of block falls and plastic deformation during tunnel excavation.

The objective of this thesis is evaluation and interpretation of various methods to assess block/wedge failures and plastic deformation along HRT at KL-III HEP. It is necessary to have a clear idea about factors triggering tunnel instabilities in weak rock condition. The scope of thesis is categorized into two main sections. Block/wedge falls have been evalu- ated with kinematic limit equilibrium (KLE) method and in UnWedge 4.0. In UnWedge, both deterministic and probabilistic methods have been applied to access the instability.

For plastic deformation, empirical methods by Singh (1992), Goel (1995) and Q-system (2000) are used. Semi-empirical methods by Jethwa (1984), Hoek and Marinos (2000) and Panthi and Shrestha (2018) have been used. Squeezing analysis and support pressure investigation have been done using the analytical method called Convergence confinement method by Carranza-Torres and Fairhurst (2000) and with improvements by Vlachopoulos and Diederichs (2009). An uncertainty analysis to access rock mass quality and plastic deformation have been included using Octave 5.1. Numerical investigation using sophis- ticated 2D and 3D finite element softwares,viz. Rocscience²& Rocscience³are included in the thesis.

Weak, fractured rock mass with high in-situ stress is the main reason for plastic deformation. In addition to that, presence of active thrust along HRT has increased the degree of wedge failures. The above mentioned methods use stress condition, rock mass quality and strength to evaluate instability. Results from analysis along HRT reveal the fact that tunneling in weak and fragile rock mass is a complicated task involving several tunnel instabilities. Thus, tunneling in such condition requires good planning and considerations.

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Chapter 1 Introduction

1.1 Background of study

Nepal is a landlocked country and it occupies the north-central position in South Asia. It is geographically, sandwiched between China in the north and India in the south. It is located in central part of the 2,500 km long Himalayan arc and covers one third of its length.

With an area of 147,516 km², it is bounded by latitude 26^◦22’ and 30^◦27’ and average east-west axis of 885 km and north south of 193 km (GoN, 2020). Geographically, major part of Nepal (∼83%) falls within mountainous region and remaining 17% is covered by alluvial plains of the Gangetic basin (Upreti, 1999). Being a mountainous country, Nepal has high potential for hydro-power development, since there is steep terrain and fast flow- ing rivers originating from glaciers which fulfill the basic requirements.

Himalaya has a very young and fragile geological formation so the construction of underground structure in this area is a challenging task. Due to active tectonic movement and dynamic monsoon, rock mass in this region is relatively weak and highly deformed, schistose, weathered and altered. The major tectonic thrust faults such as Main Central Thrust (MCT) and Main Boundary Thrust (MBT) have significant influence on the high degree of shearing and fracturing to rock mass. Predicting rock mass quality, analyzing stress induced problems, in particular tunnel block fall and squeezing, often have been found extremely difficult. The impact of continuous collision of continents till now results several thrusts and faults in Himalaya (Panthi, 2006).

Based on these faults and thrusts as well as rock type and age, Nepal Himalaya can be divided into five east-west trending major tectonic zones. They are: Terai, Siwalik, Lesser himalaya, Higher himalaya and Tibetan Tethys zones. Mainly, there are two types of stability problems. They are block fall & squeezing in weak and deformable rocks and rock burst & spalling in strong and brittle rock. In Himalayas of Nepal, tunnel squeezing is a common phenomenon as the fault zone weak rocks like mud stone, slate, phyllite, schist and highly schistose gneiss, that compose the mountains are not capable of withstand- ing high stress (Panthi, 2006). Many hydropower tunnels like Kaligandaki-A, Middle

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Marsyangdi, Modi and Khimti respectively have considerable amount of squeezing and block falls while tunneling (Panthi, 2006). These tunnels lies in Lesser Himalaya and have high overburden pressure with weak geology.

Kulekhani-III Hydroelectric Project (KL-III HEP) is one of such projects, situated in central lesser himalaya in Makawanpur district of Nepal. The headrace tunnel passes through major thrust zones and there has been reported incidents of major block falls, wedge failures and plastic deformation along tunnel system. Therefore, it is interesting to study various immediate and long term instability issues and their possible causes along this headrace tunnel.

1.2 Objective and Scope

Main objectives for the study are:

• Evaluate and interpret block and wedge failures in headrace tunnel at KL-III HEP.

• Evaluate and interpret plastic deformation along headrace tunnel at KL-III HEP.

The scope of the project involves:

• Review existing theory on the stability issues in underground excavation with focus on the block falls and plastic deformation.

• Briefly describe about KL-III HEP and discuss about the extent of engineering geological investigations carried out at the project.

• Assess predicted and actual rock mass conditions as well as rock support use along the tunnel alignment.

• Evaluate block falls and plastic deformation using prevailing rock engineering theory. While evaluating block falls, seismic load shall be considered.

• Analyse block falls using wedge unwedge program considering seismic load. Anal- yse plastic deformation using numerical modelling for the selected segments of the headrace tunnel.

• Compare and discuss the analysis results from analytical and numerical approaches.

1.3 Methodology of study

The following methodology has been applied during study:

1. Literature review:

(a) Review of the geology of the Himalaya of Nepal regarding rock type, stress induced, weathering effect on rock mass.

(b) Review of stability issues in underground tunnels in weak rock mass.

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(c) Review of existing empirical, semi-empirical and analytical methods to evaluate block falls and plastic deformation.

(d) Review on numerical investigation methodology.

2. Study of Kulekhani-III Hydroelectric project:

(a) Overview of layout of project with focus on headrace tunnel of project.

(b) Evaluate engineering geological condition and rock mass properties along headrace tunnel.

(c) Evaluate and interpret block fall and plastic deformation around critical locations in HRT.

3. Instability analysis:

Based on available data, the analysis of block fall and plastic deformation has been done using different methods as listed below:

(a) Limit equilibrium analysis:

(b) Empirical methods: Singh et al. (1992) method; Goel et al. (1995) method and Q-system.

(c) Semi-analytical methods: Jethwa et al. (1984) method; Hoek and Marinos (2000) method and Panthi and Shrestha (2018) method.

(d) Analytical method: Convergence confinement method by Carranza-Torres and Fairhurst (2000).

(e) Uncertainty analysis:

4. Numerical Investigation:

Wedge stability in HRT is analysed with UnWedge 4.0. Rock squeezing is analyzed using commercially available finite element software, RocScience v.9.0. Lab tested rock mass parameters by NEA (1997) are checked with available back calculations.

Detailed study is done with verified rock mass parameters.

5. Comparison and Interpretation of results:

Results from all above methods are compared with each other and necessary inter- pretations are made.

1.4 Limitations

The main limitation for the study is reliability of input geological variables. The study is based on geological data, reports and documents that author had accessed via NEA. So, data are not first hand. Identification of failure modes in tunnel is a difficult task due to limited access to good and clear pictures. There has not been any study done regarding horizontal stress condition around the project site. Tectonic stress measurements have been adopted from the nearest hydropower tunnel with similar geological conditions because of no actual measurements carried out in project.

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Limited tests and measurements have been major limitations for the study. NEA has carried out in-situ rock tests and laboratory measurements of the rock specimens to find out the rock mass properties. Author has done the level best to screen data and use the most reliable one. Principle of back calculation from measurements have been used to validate the parameters to be used for analyses.

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Chapter 2 Rock and rock mass properties

2.1 Introduction

There is a crucial difference between rock and rock mass. Rock is a naturally composed aggregate of minerals whose properties depend on mineral composition, orientation, size, shape and binding forces between minerals (Nilsen and Thidemann, 1993). Rock mass, in other hand, is a structural construction material that composes of intact rock, joints and discontinuities. According to Panthi (2006), rock mass is a heterogeneous medium that characterizes two main features, namely, rock mass quality and mechanical proprieties in rock mass, which have been illustrated by Figure 2.1.

Tunnel Stability

Mechanical processes Rock mass quality

Rockstresses Groundwater Rockmassstrength Strengthanisotrophy

Rockmassdeformability Weathering&Alteration

Discontinuity

Figure 2.1:Factors influencing tunnel stability. Redrawn after (Panthi, 2006).

Rock mass strength, rock mass deformability, strength anisotrophy, faults and discontinuities and weathering affect rock mass quality. Similarly, rock stresses and ground water affect mechanical processes. In addition to these, there are some other factors such as size, shape and orientation of tunnel which affect stability of tunnel. This chapter gives an

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overview on these rock mass properties which influence the overall stability of tunnel.

2.2 Rock mass strength and deformability

The uni-axial compressive strength (σ_ci) and intact deformation modulus (E_ci) of rock samples are tested at in-situ condition or at lab (Nilsen and Palmstr¨om, 2000). The rock mass strength (σ_cm) and deformation modulus (E_rm) are often estimated from classification systems indirectly because of operational difficulties and time consuming in-situ rock tests (Palmstr¨om and Singh, 2001). Tests carried out by NEA (2003) at headrace tunnel have been used in analysis.

2.2.1 Intact rock strength

An intact rock is strong and homogeneous with few discontinuities and stronger than the rock mass (Bieniawaski and Van Heerden, 1975). Intact rock strength and deformability are determined from lab test or in-situ tests (Nilsen and Thidemann, 1993). Uni-axial compressive strength (σ_ci) test is the most common method to test mechanical characteristics of rock where intact rock specimen cylinder is loaded till failure. Value ofσ_ciis useful in calculating the rock mass strength as suggested in section 2.2.2. Factors that influence intact strength will thus influence rock mass strength too. Some of these factors are: Size of specimen, strength anisotrophy, water effect and weathering & alteration which are described below.

Scale size effect

Intact rock specimen is strong and homogeneous with very few discontinuities. Therefore the strength of this specimen doesn’t represent strength and deformability of actual rock mass. There is significant scale effect. With increase in size of specimen,σcidecreases.

Uniaxial compressive strength test performed on a smaller size specimen will thus have higher strength (Hoek, 2007), which is indicated in Figure 2.2a. Size effect is also affected by metamorphism. Crystalline unweathered rocks have less size effect but highly schistose folliated and weathered rock mass have high size effect (Panthi, 2006).

Strength anisotropy

Mineral grains orientation and directional stress history are the major causes of strength anisotropy in rock mass. Anistoropy is maximum in sedimentary rocks (layering and bedding planes) and metamorphic rocks where minerals are mostly in form of parallel and weak layers inform of schistocity and folliation (Nilsen and Palmstr¨om, 2000). The maximum strength of intact rock is obtained when schistocity plane is normal to loading direction and the minimum strength is obtained when the angle is∼30^o, as illustrated in Figure 2.2b (Panthi, 2006). Degree of anisotrophy is determined through point load test. Initially, point load test is normal to folliation and then parallel. The ratio of the two strength gives degree of anisotophy.

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Figure 2.2:Curve showing influence of specimen size of rock core on uniaxial compressive strength of intact rock (Nilsen and Palmstr¨om, 2000) (left). Variation of uniaxial compressive strength of different rock type at different schistocity plane angle (Panthi, 2006) (right).

Weathering effect

Weathering is the physical breakdown/disintegration and chemical decomposition if rock at/ near the earth surface (ISRM, 1978). Degree of weathering decreases the rock mass strength and Panthi (2006) has quantified the extent of lowering of rock strength with severity of weathering grade which has been illustrated in Figure 2.3.

Figure 2.3: Variation of uniaxial compressive strength of different rock type (left) and strength reduction in % as the function of weathering grade (Panthi, 2006).

According to Nilsen and Palmstr¨om (2000) mechanical disintegration involves rock losing its coherence causing opening of joints, grain boundaries and mineral grains fracturing.

Similarly, chemical disintegration involves rock decaying resulting discoloration, decomposition of silicate and leaching of calcite (Nilsen and Palmstr¨om, 2000). ISRM (1978) has classified weathering grade into six different categories which are shown in Table 2.1.

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Table 2.1:Classification of weathering grade according to ISRM(1978) (Panthi, 2006).

2.2.2 Rock mass strength

According to Panthi (2006), rock mass strength (σ_cm) is the ability to withstand stress and deformations and is influenced by discontinuities, foliation and their orientation in which the strength is assessed. As compared to intact rock strength, rock mass strength is difficult to estimate through field work or laboratory test so many researchers have therefore, suggested empirical relationships for the estimation ofσ_cm. Almost all methods discussed here include intact rock strengthσ_cilinked with the rock mass characterization parameter such as Q-value or Rock mass rating (RMR) to estimateσ_cm(see Table 2.2).

Table 2.2:Empirical relationships for estimation of rock mass strength (Panthi, 2006).

Proposed by Empirical relationship

Bieniawaski (1993) σcm=σci×exp(^{RM R−100}_18.75 )

Hoek et al. (2002) σ_cm=σ_ci×exp(^GSI−100_18.75 )^a =σ_ci×exp(^{RM R−105}₉ )^a Barton (2002) σcm= 5γ×(^σ₁₀₀^ci ×Q)¹³ = 5γ×(^σ₁₀₀^ci ×10^{RM R−50}¹⁵ )

1 3

Panthi (2006) σcm= ^σ₆₀^1.5^ci

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Among these rock mass strength estimation methods, the first three methods relating RMR

&σ_civalues toσ_cmhave a drawback when are being used for schistose and highly foliated rock masses. The reduced strength of discontinuous rock is double accounted for while findingσ_ciin lab and determining RMR, Q. GSI is geological strength index (see Appendix A.2). The relation proposed by Panthi (2006), relatesσ_cmonly withσ_civalues.

This relation is the power plot ofσ_civersusσ_cmbased on top three relations in Table 2.2.

2.2.3 Rock mass deformability

According to ISRM (1975), the modulus of elasticity of intact rock (E_ci) is the ratio of stress to corresponding strain within proportionality limit. Rock mass deformability (E_rm) is the ratio of stress to corresponding strain during loading of rock mass which includes both elastic and inelastic performance (wd) as shown in Figure 2.4 (right). Elasticity modulus of rock mass (Em) is the ratio of stress to corresponding strain during loading of rock mass which involves only elastic performance (we), see Figure 2.4 (right). There are two distinct methods to estimate (Erm), viz. in-situ measurements such as plate loading test (PLT), Plate jacking test (PJT) and Radial jacking test (Goodman jack test).

In-situ deformation tests are expensive and difficult to conduct (Palmstr¨om and Singh, 2001). Therefore, many researchers have suggested empirical equations to estimate Erm

indirectly based on observations of relevant rock mass parameters which are given in Table 2.3. Since the project under study has done PLT, see Figure 2.4 (left) at some sections of headrace tunnel, the results will be compared with indirect estimations for valida- tion.

Figure 2.4: Principle of plate loading test (PLT) (left) and typical stress versus deformation curve recorded in a deformability test of a rock mass (right) (Palmstr¨om and Singh, 2001).

In Table 2.3, Ermis modulus of deformation in GPa,Eciis modulus of elasticity in GPa andDis the disturbance factor (see Appendix A.3).

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Table 2.3:Empirical relationships for estimation of rock mass deformability (Panthi, 2006).

Proposed by Empirical relationship

Bieniawaski (1989) Erm= 2RM R−100

Hoek et al. (2002) Erm= (1−0.5×D)×pσci

100×10⁽^GSI−10⁴⁰ ⁾

Barton (2002) Erm = 10×Q

1

c3 = 10×(^σ₁₀₀^ci ×Q)¹³

Hoek and Diederichs (2006) Erm=Eci×

0.02 + ¹⁻^D²

1+e^60+15D−GSI11

Panthi (2006) Erm=Eci×^σ₆₀^ci^1.5

2.3 Discontinuities

According to Nilsen and Palmstr¨om (2000), discontinuities are the weakness planes along the rock mass which have zero to nearly zero tensile strength. Most often, they represent joints, weakness planes, fault zones, bedding surface and folliation surfaces. Discontinuity is a governing factor that determines mechanical properties of a rock mass in engineering geology. Figure 2.5 shows the distribution of major types of discontinuities (Grimstad and Barton, 1993).

Figure 2.5:Size distribution of different types of discontinuities according to Grimstad and Barton (1993).

2.3.1 Joints in rock mass

Joints are defined as the fracture or the break of natural continuity of rock mass which has no noticeable shift (Mandl, 2005). A joint set consists of number of parallel joints.

When a joint set intersects with another, a joint system is formed. Random joints without

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any specific pattern can also be present in rock mass. Figure 2.6 illustrates major joint characteristics required to be identified in geological mapping. Based on the origin and process of formation, Nilsen and Palmstr¨om (2000) have categorized joints as follows:

Tectonic joint:Joints formed in a tensile stress field after tectonic disturbances.

Exfoliation joint:Parallel surface joints formed due to heating and thawing of rock mass.

Bedding stress:Joints along bedding planes in sedimentary rocks.

Folliation stress:Joints along folliatin planes in metamorphic rocks.

Sheet stress:Surface joints in igneous rocks formed due to rock mass unloading by erosion.

Figure 2.6:Joint characteristics in a rock mass (Hudson and Harrison, 1997).

Following are the most critical joint characteristics:

Alteration & Filling

Alteration & filling of discontinuities in a rock mass is due to processes such as weathering, rock mass shearing and hydro-thermal alteration. When the surface of rock mass is weathered and altered to toughness, this changes the joint wall strength. Rock mass having joints without any filling are stable while the joint with soft and thick filling lowers the friction and hence the stability (Palmstrøm and Stille, 2015).

Size of block

Joints in rock mass forms block whose size can vary from each other and is dependent on joint spacing, joint sets and persistence of joint (ISRM, 1978). Rock mass with larger blocks tends to deform less and have efficient arch effect and interlockablility. But rock mass with smaller block has lesser arch effect and interlockability.

Surface roughness

According to ISRM (1978), surface roughness is delineated as the large scale waviness

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and small scale smoothness (unevenness). In context of rock mass having discontinuities with high strength or low stress level, surface roughness causes dilation during the shear movement. The chart shown in Figure 2.7 (left) estimates relationship betweenJrfrom Q- system andJRCn. Figure 2.7 (right) as described by Barton (1988), shows categorization of different roughness profiles.

Figure 2.7: Relationship betweenJaand residual friction angle (φr) (left) andJrandJRCn (right) (Barton, 1988).

Persistence

ISRM (1978) describes persistence in rock mass as the discontinuity length within a plane.

In other word, it is the length, size and areal extent in a plane. Persistence can be either continuous type which terminate in other joints or discontinuous type which ends in massive rock mass (Palmstrøm and Stille, 2015).

2.3.2 Shear strength of Joints

According to Barton and Bandis (1990) shear strength of rock sample can be estimated using equation (2.1).

τ=σn×tan

J RC×logJ CS σn

+φr

(2.1) where, JRCis joint roughness coefficient which measures depth of groove in the rock mass. JCSis Joint compression strength and is measured with a Schmidt hammer andφ_r is the residual friction angle and depends on the tilting angle. Barton and Bandis (1990) suggested that the residual friction angle is calculated as equation (2.2).

φr= (φb−20) + 20r

R (2.2)

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where,φ_bis the basic angle of friction,ris Schmidt rebound on wet joint surface andRis the rebound value on dry surface.

2.3.3 Joint stiffness

According to Barton (1972), joint stiffness can be related to property of the intact rock, rock mass and joint spacing. As per Goodman (1989), the normal joint stiffness (K_n) and the shear joint stiffness (K_s) can be expressed as equations (2.3) and (2.4) respectively.

K_n= E_ci.E_rm

L(E_ci−E_rm) (2.3)

K_s= Gci.Grm

L(G_ci−G_rm) (2.4)

where,

Eciis Intact rock modulus,Ermis rock mass modulus,Gciis Intact shear modulus,Grmis rock mass shear modulus andLis Joint spacing.

Shear modulus (G) is related to rock modulus (E) and Poisson ratio (ν) and is expressed as equation (2.5) (Panthi and Shrestha, 2018).

G= E

2(1 +ν) (2.5)

2.3.4 Faults and weakness zones

According to Nilsen and Palmstr¨om (2000), weakness zones are the part of the rock mass where mechanical properties are much lower than the surrounding rock mass. Weakness zone formed by tectonic movements are called as faults/ faulting zone or fracture zone. The movement is mostly due to the shearing of rock mass. The zone can vary significantly in distribution based on the extent of jointing, filling and weathering (Nilsen and Palmstr¨om, 2000). Bed or layer of weak minerals such as clay, mica, talc and graphite make the rock mass much weaker. For instance, in mica schist, presence of mica significantly reduces the strength of rock mass. Figure 2.8 shows some common types of weakness zones.

Figure 2.8:Different types of weakness zones. Shaded region shows altered rock and block region are filling (Palmstrøm and Stille, 2015).

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Most often, the weakness zone composes of a central core surrounded by a transition zone.

Central core is highly fractured zone and is composed of an altered rock mass with clay fillings. In other hand, transition zone composes fractures with decreased frequency and size away from the central core (Panthi, 2006).

2.4 Failure criteria

In an engineering geology, a failure of rock mass indicates the loss of integrity leading to collapse of load carrying capacity of rock mass. Many theories have been developed to ex- plain prediction of failure in rock mass. The classical theoretical failure criterion include Tresca criteria, Mohr-Coulomb criteria, Drucker-Prager criteria and Griffith criteria. The most commonly used criteria is Mohr Coulomb criteria which is based on maximum effective shear stress (Hudson and Harrison, 1997). Since, theoretical failure criteria doesn’t actually reflect the actual nature of failure, Generalized Hoek and Brown failure criterion which is based on experimental failure plot inσ₁-σ₃plane is widely used (Hudson and Harrison, 1997).

2.4.1 Mohr Coulomb failure criterion

Mohr Coulomb failure criterion is a linear criterion useful to assess stability of tunnels in isotropic, unjointed and elastic rock mass. Mohr-Coulomb failure criteria shows the relationship between shear and normal stress at failure. Most often, it is termed as inner friction criterion since an internal friction for material expressed with friction angleφis used.

Mohr Coulomb failure criterion consisting of a liner Mohr envelope that touches different Mohr circles at different confining stresses can be defined as illustrated in Figure 2.9 (left) for a rock core specimen loaded until failure as shown in Figure 2.9 (right).

Figure 2.9:Mohr-Coulomb failure criteria (Hudson and Harrison, 1997).

The failure envelope is represented with equation (2.6).

τp=c+σntanφ (2.6)

where,τis shear stress,σnis the normal stress and c is cohesion.

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The plane of maximum shear stress is inclined at angle of 45^◦to horizontal and is given as equation (2.7).

τ_max=σ₁−σ₃

2 (2.7)

According to Goodman (1989), tensile cutoff (To) is usually used in the criteria since the criteria is used for compressive stress. Mohr Coulomb criteria is recommended for high confining pressure when the failure occurs through shear planes (Hudson and Harrison, 1997).

2.4.2 Generalized Hoek-Brown failure criterion

Generalized Hoek-Brown failure criterion is a nonlinear failure criterion to assess stability of tunnel in jointed and schistose rock mass. The criterion shows the empirical relationship based on fitting of parabolic curves in triaxial test data (Nilsen and Thidemann, 1993).

According to Hoek and Brown (1980), the original Hoek and Brown failure criterion for intact rock is given as equation (2.8).

σ₁=σ₃+σ_ci r

mσ3

σ_ci+s (2.8)

where, σ₁ andσ₃ are major and minor principal stresses. σ_ci is Uniaxial compressive strength of intact rock. m and s are constants which depend upon rock properties.

The criteria has been updated based on the experiences and practical situations (Nilsen and Thidemann, 1993). Hoek et al. (2002a) suggested that for the jointed rock mass the failure criteria is defined by equation (2.9).

σ⁰₁=σ₃⁰ +σci mσ₃⁰ σci

+s

!^a

(2.9) where,σ1’andσ3’are major and minor effective principal stresses.mbis a Hoek & Brown constant that depends onmiand s & a are rock mass constants.mb, s & a are calculated as equations (2.10), (2.11) and (2.12) .

m_b=m_i×exp

GSI−100 28−14D

(2.10)

s= exp

GSI−100 28−14D

(2.11)

a= 1 2+1

6

e^−GSI/15−e^−20/3

(2.12) where,m_iis intact rock constant, GSI is Geological strength index andDis the disturbance factor (0 to 1) depending on disturbance in rock mass due to blasting and stress relaxation (Hoek et al., 2002a).

Based on illustration in Figure 2.10, Hoek and Brown failure criterion is best suited for intact rock or rock mass having closely spaced joints with similar characteristics.

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Figure 2.10:Selection of failure criteria based on rock mass condition. Modified from (Hoek et al., 2002a).

2.4.3 Post failure characteristics

Based on quality of rock mass, they have different post peak behavior. According to Hoek (2007), a massive brittle rock mass with high Q value will loose strength quickly when the maximum strength is exceeded. Significant dilation is often a result. Thus, it is best delineated by an elastic brittle plastic material Figure 2.11 (a). A medium-quality rock masses will result in stress-strain evolution shown with a strain softening effect Figure 2.11 (b). A poor quality rock masses show nearly perfect plastic behavior Figure 2.11 (c).

Figure 2.11:Post peak behaviour in different rock masses (a) Good quality (b) Average quality (c) Poor quality rock masses (Hoek, 2007).

Cai et al. (2007) have quantified residual strength of rock mass based on GSI system using block volume and joint conditions. According to Cai et al. (2007), the post failure characteristics of rock mass is dependent on new failure surfaces and blocks interlocking

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which are quantified with equation (2.13) and illustrated with Figure 2.12.

GSI_r=GSI×e^−0.0134GSI (2.13) After finding out value of residualGSI, residual strength parameters are calculated using

Figure 2.12: Relationship between GSI and GSIr. The right dotted line represents the result of Russo et al. (1998), where GSIris estimated to be 36% of GSI (Cai et al., 2007).

Generalized Hoek and Brown failure criteria given with equation (2.14), equation (2.15) and equation (2.16).

m_r=m_i×exp

GSI_r−100 28

(2.14)

sr=exp

GSIr−100 9

(2.15)

a_r= 0.5 + 0.167

e⁻^GSIr¹⁵ −e⁻²⁰³

(2.16) Determination of residual GSIs are shown in Appendix A.4. Peak values of GSI are low- ered to the corresponding residual state with respect to combination of joint surface conditions and block volume degradation. The values of intact rock parameters, i.eσciandmi

remain unchanged during residual values calculation. Since the rock masses are in dam- aged, residual state so value of D = 0 is used for residual strength parameter determination (Cai et al., 2007).

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Chapter 3 Stresses and instabilities in tunnel

3.1 Rock stresses

When induced stress around tunnel is greater than critical stress (strength) for rock mass, tunnel fails. Stress in a tunnel are defined by two major features: Stress situation before excavation (virgin stress) and tunnel geometry (Nilsen and Thidemann, 1993). Excavation in tunnel alters the stress state in rock mass surrounding. Stress situation is thus, function of initial stress state and stresses induced due to excavation. So, in-situ stress measurements are important during stability assessment of a tunnel.

3.1.1 In-situ stress in rock mass

According to Nilsen and Thidemann (1993), in-situ stresses are the resultant of following components:

Gravitational stress:Rock stress originated due to gravity alone.

Topographic stress:When surface is not horizontal, topography affects stress.

Tectonic stress:Stress produced due to tectonic movement.

Residual stress: Remnant stress which has been locked into rock material during earlier stage of geological history.

Principal stresses are the representation of resultant of stresses at any point in the rock mass. A principal stress is the stress in a principal plane, which is acted upon by, only normal stresses and no shear stresses. There are 3 principal stresses namely, Major principal stress (σ1), Intermediate principal stress (σ2) and Minor principal stress (σ3). Nilsen and Palmstr¨om (2000) have given a way to measure these stresses. Due to overburden, stress gets induced in two directions, namely vertical and horizontal directions. At a depthzbe- low the rock surface, and with the horizontal surface assumption, the vertical gravitational stress is given as equation (3.1).

σv =σz=γrH (3.1)

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where, γ_r is specific gravity of rock mass in MN/m³. There have been many site measurements of vertical stresses from tunnels and minings around the world plotted against depth. As indicated by Carranza-Torres and Fairhurst (2000), after (Hoek and Brown, 1980), Figure 3.1a shows the relationship producing a straight line with a slope of 0.027 which is fairly correct since the value is mean specific gravity of rock. The values however, can have significant deviations near the surface or at considerable depth because of lim- its of measuring equipments at shallow depths and presence of residual stresses at great depth (Nilsen and Palmstr¨om, 2000). With reference to Panthi (2006), horizontal stress is calculated as the sum of effect due to gravity and tectonic condition and is given as equation (3.2).

σ_H =σ_tect+ ν

1−ν.σ_v (3.2)

According to Hoek (2007), the value of stress factor (k) vary significantly due to both topography and the tectonic movements which is illustrated in Figure 3.1b. These variation, therefore highlights the importance of in-situ stress measurement before undertaking any underground excavation. Most often tectonic stress do not align normal and parallel to

(a) (b)

Figure 3.1:Vertical stress as the function of depth (z) (left) and variation of stress factor with reference to depth from surface (right) (Carranza-Torres and Fairhurst, 2000), after (Hoek and Brown, 1980).

the HRT alignment, therefore necessary resolving in equivalent in-plane and out-of plane stresses are done using equations (3.3) and (3.4).

σin-plane=σtect.cosβ+ ν

1-ν.σvertical (3.3)

σout-plane=σtect.sinβ+ ν

1-ν.σvertical (3.4)

where,βis angle made by tectonic stress with the equivalent in-plane stress.

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3.1.2 Stress distribution around tunnel

Excavation of underground opening in a rock mass causes in-situ or virgin stress to redis- tribute around opening and changes in stress magnitude is seen (Nilsen and Palmstr¨om, 2000). This is illustrated in Figure 3.2 (left). In case of circular tunnel with elastic material and isostatic stress state, i.eσ_h andσ_vare equal to each other,Kirsch solution, as described by equation (3.5) and equation (3.6) are applicable to calculate tangential stress (σθ) and radial stress (σ_r) (Panthi, 2006).

For an elastic material,σθwith the magnitude of two times principal stress (σ) is induced all around the periphery. Stress lowers down asymptotically and normalizes to a constant level equal toσat distance corresponding to approximately half of tunnel width (Nilsen and Palmstr¨om, 2000) as illustrated in Figure 3.2 (right: solid line).

σθ=σ

1 + r² R²

(3.5)

σr=σ

1− r² R²

(3.6) In case of non-isostatic stress state,Kirsch solutionstates thatσθreaches its maximum

Figure 3.2:Distribution of stress around rock mass with a circular opening (left) and tangential &

radial stress distribution in elastic and plastic sates (right) (Panthi, 2006).

value σθ,max where σ1 direction is a tangent to the tunnel contour and it’s minimum value σ_θ,min, where σ₃ direction is a tangent and are described with equation (3.7) and equation (3.8).

σθ,max = 3σ1−σ3 (3.7)

σθ,min= 3σ3−σ1 (3.8)

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If the stresses are very anisotopic, minimum tangential stresses can even be negative, i.e tensile. For a weak rock mass, when rock mass strength is lower than the tangential stress, then a plastic zone is seen around the opening as shown in Figure 3.2 (right) (Shrestha, 2005). Since, plastic zone is unable to take any more stress, stress peak shifts out from rock opening contour, see Figure 3.2 (right: dotted line). Detail on solution of stresses and displacements from plasticity theory by Goodman (1989) can be referred for further analysis.

According to Xiao et al. (2019), in an elasto-plastic rock mass, the radial and tangential stresses in elastic zone are similar as in Hoek & Brown but the radial distance and stresses of elasto-plastic boundary are different. Under Mohr-Coulomb media, the radius of yield zone is calculated with equation (3.9).

re=ri[1−sinφ)(σv+c/tanφ](1−sinφ)/2 sinφ

(3.9) The radial stress in a radius of yield zone is given as equation (3.10).

σ_re= (1−sinφ) [σ_v−c/tan((φ/2) + 45^◦)] (3.10) For a non-circular opening, Hoek and Brown (1980) have carried out a detailed stress analysis using a boundary element technique and have suggested a practical method to estimate the tangential roof stress (σ_θr) and tangential wall stress (σ_θw) which as estimated with equation (3.11) and equation (3.12) respectively.

σθr= (A×k−1)σz (3.11)

σθw= (B−k)σz (3.12)

where, A & B are the wall factors for tunnel shapes as illustrated in Figure 3.3.

Figure 3.3:Wall factors for different tunnel shapes as suggested by Hoek and Brown (1980).

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3.2 Tunnel stability issues

A rock mass has rock blocks and fragments with discontinuities (Nilsen and Palmstr¨om, 2000). The properties of rock mass is determined by the arrangement of joints and the filling in them. Degree of tunnel instability is directly dependent on the rock mass condition, jointing, weakness zones and stress orientation. According to Hudson and Harrison (1997), tunnel instabilities can be broadly classified into two categories, namely structurally controlled failure and stress induced failure.

3.2.1 Structurally controlled failure

The presence of preexisting blocks in rock mass fall and/or slide produces instability around tunnel which are called as Structurally controlled failure. Wedges are formed as a result of intersection of three major plane of weaknesses and excavation boundary as the fourth boundary. The wedges detach out as a result of poor restrains from surrounding rock mass. The orientation of joint sets, shape of tunnel and rock mass condition including weathering extent and friction influence structurally controlled failures. KL-III HRT has experienced serious wedge fall and block fall. Therefore, this issue shall be dealt in depth in subsequent chapters.

3.2.2 Stress induced failure

When induced stress is greater than the rock mass strength then stress induced failures are seen around tunnel (Panthi, 2006). The failure can be in form of rock spalling/ bursting in strong, massive and brittle rock mass or rock squeezing/ plastic deformation in weak, deformable and ductile rock mass. Hoek (2007) has defined rock bursting as the ‘explo- sive failure of rock due to high concentration of stress around tunnel’. There are three distinct mechanisms of failure,viz. rock bulking from fracturing, rock ejection from seismic energy transfer and rockfall from earthquake. The stress induced failure propagation in brittle rock mass is illustrated with Figure 3.4. Since HRT at KL-III HEP doesn’t have any kind of rock spalling problem so this won’t be dealt further.

Figure 3.4: Stress induced failure propagation in brittle rock mass a) Non violent spalling b)Rock burst due to buckling c) Spalling intersection with structure d) Dilationlal yield as suggested by Diederichs (2014).

Weak rock masses are characterized by gradual inward movement of rock mass towards tunnel opening which is termed as rock squeezing. Aydan et al. (1993) has provided three

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possible forms of failures in squeezing ground conditions as shown in Figure 3.5. Com-

Figure 3.5: Stress induced failure in squeezing rock mass a) Complete shear failure b) Buckling failure c) Tensile splitting shearing and sliding as suggested by Aydan et al. (1993).

plete failure indicates complete shearing of rock mass and is seen in rock mass with widely spaced discontinuities. Buckling failure is seen in thinly bedded ductile rock masses.

Shearing and sliding failure involves sliding along bedding plane and intact rock shearing and is seen in thick bedded sedimentary rock (Aydan et al., 1993). HRT at KL-III has experienced squeezing problem at some specific locations so this failure problem will be dealt in detail.

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Chapter 4 Instability evaluation methodologies

This chapter deals with different methodologies to evaluate tunnel instability along HRT at Kulekhani-III hydroelectric project. Tunnel instabilities include block and wedge failures along shear zones including sections passing Mahabharat thrust and plastic deformation in weak rock masses like phyllite with overburden as high as 315 m. This research aims at utilizing as many methods as available to identify the cause of instability and recommends the best way to overcome similar kind of problems in other projects with similar site conditions. As for block and wedge fall, traditional deterministic method and numerical modeling with UnWedge 4.0 have been utilized. In case of plastic deformation, methods such as empirical, semi-empirical, analytical method like Convergence Confinement Method (CCM) by Carranza-Torres and Fairhurst (2000) and probabilistic uncertainty analysis are used. In addition to that numerical modeling with 2 & 3 dimensional finite element softwares, Rocscience 9.0 have been used to analyze stress induced instability.

4.1 Block and wedge fall

When a tunnel is excavated in a highly foliated and jointed rock mass, the most common tunnel instability seen are block and wedge failures. There are three weakness plane, for instance, bedding plane or discontinuities in the rock mass along excavating surface of tunnel for wedge formation as shown in Figure 4.1 (Hoek, 2007). Block fall also follows similar pattern where failures are in form of blocks. The weight wedge or block is major driving force and the orientation of discontinuities with respect to tunnel alignment determines shape and volume of failures (Nilsen and Palmstr¨om, 2000).

When block or wedge fall occur, the interlocking capacity of rock mass gets disturbed and thus, more failure start occurring. This phenomenon continues till a full of material (natural arc) in rock mass is built up and thereby restricting further collapse of tunnel. The phenomenon is termed asKey blockby Goodman (1989).

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Figure 4.1:Roof wedge fall (left) and sidewall wedge fall (right) occurring by intersection of three weakness planes (Hoek, 2007).

4.2 Evaluation methodologies of block fall & wedge fall

4.2.1 Kinematic limit equilibrium (KLE) analysis

Limit equilibrium method is one of the most common block fall evaluation methods used in engineering geology to estimate stability of wedge and block around tunnel opening.

To make it simple, all forces acting in the wedge or the block are identified and then cat- egorised into two different type of forces, namely resisting force (Fr) and driving force (Fd). Resisting forces are the stabilizing forces such as frictional resistance of joint de- scribing its shear resistance. In other hand, driving forces are the destabilizing forces such as weight of the block/ wedge, seismic condition or water pressure which disturb the stability of block and cause movement.

A simplified condition of block and wedge fall is shown in Figure 4.2. Weight of block acts vertically downward. Normal force acts in sliding plane and resisting force act op- posite to direction of sliding. The stabilizing or resisting force is given as equation (4.1) (Barton and Choubey, 1977).

W N

N S

S

�

(a)

S

h

l l

W

N N

S

W

N

N S

S

(b)

Figure 4.2:Illustration of forces for a simplified situation for block and wedge failure.

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Fs=τ×A (4.1) For a limit equilibrium analysis, shear strength of joint is given with equation (4.2) (Barton and Choubey, 1977).

τ =σntanφa (4.2)

where, φ_a is the active friction angle including all shear resistance with apparent cohesion. Here, the shear strength of the joints over the area of slidingτ is important property resisting block from failure. Weakening of joint is represented by reduction in joint compression strength, roughness coefficient and angle of internal friction of the rock mass.

Shear strength of rock sample is difficult to test due to scale effects and getting repre- sentative sample therefore Barton and Bandis (1990) equation is generally used which is given by equation (2.1). Since there is no prominent water source above tunnel so tunnel is mostly dry and water pressure isn’t that significant. The driving force in block is given with equation (4.3) (Barton and Choubey, 1977).

Fd=W +Ncosϕ−Ncosϕ=W (4.3)

In case of wedge, the forces acting on it is illustrated in Figure 4.2b. The edge length (l) is calculated with equation (4.4).

l=L/cosϕ (4.4)

Similarly, area of edge is calculated as product of edge length and depth. Volume of wedge is thus calculated with equation (4.5).

V =h²d∗tanϕ (4.5)

The driving forces for wedge failure is given with equation (4.6).

F_d = (W + 2N.sinϕ) (4.6)

Maximum tangential stress is seen at the crown of the tunnel as shown in Figure 8.1a.

Radial and axial stresses are of less significance and may be neglected over here. Normal stress on wedge edge due to this tangential stress is given with equation (4.7).

N =σ_nθ.A (4.7)

Factor of safety against block/ wedge fall is given with equation (4.8).

F S= Stabilizingf orce

Destabilizingf orce (4.8)

4.2.2 UnWedge analysis

UnWedge is a three dimensional numerical tool for analysing the geometry and stability of underground opening in a rock mass with structural discontinuities. Goodman and Shi’s

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block theory are the foundation to wedge stability analysis in UnWedge. Potentially un- stable wedges are modeled and support system requirements are checked around tunnel opening to calculate factor of safety.

Deterministic Analysis in UnWedge

The principle of deterministic analysis involves determination of factor of safety with the fixed values of effective parameters (Wyllie and Mah, 2004). UnWedge software has a built-in option to calculate safety factor of falling wedge under excavation and supporting of tunnel in rock mass (see Figure 4.3 (left) & Figure 4.3 (right)). Following steps are used to deal with wedge/block failure in the software.

• Determination of average dip and dip direction of discontinuity sets.

• Identification of potential wedges.

• Factor of safety calculation.

• Reinforcement calculation.

(a) (b)

Figure 4.3: Wedges formed in roof & sidewalls in an opening in a jointed rock mass (left) (Hoek, 2007). Rock bolting pattern for stabilizing roof & sidewall wedges (right) (Hoek, 2007).

Probabilistic Analysis in UnWedge

There are various uncertainties related to input parameters such as rock mass strength, the inclination and orientation of discontinuities in rock mass and internal friction of rock joints to be used in the program. Therefore, instead of using a single precise input value, a range of input values can best address the real site problem. So, probabilistic analysis basically deals with calculating safety factor distribution for each of potential wedges from which probabilities of failures are calculated. The interface allows user for pseudo- random number or random number generation. The input parameters are distributed about their mean, described by one of the distribution functions such as normal distribution and

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distribution of safety factor and probability of failure are calculated. Figure 4.4 highlights principle of probabilistic analysis in UnWedge for an underground opening.

tr

Relative frequency

Input variable (x)

y=f(x)

Relative frequency

Output variable (y) Output variable (y)

Cumulativefrequency

Opening in a rock mass

Input probability distribution model Output probability distribution

Input variables Eg: JRC, JCS, Friction angle

Output variables Eg: FS, wedge depth, wedge weight, support pressure

Figure 4.4:Illustration of principle of probabilistic analysis in UnWedge for an underground opening. Based on (Panthi and Nilsen, 2007b).

Monte Carlo method orLatin Hypercubesampling are common sampling methods. Monte Carlo simulation uses random or pseudo-random numbers for sampling probability density functions which are specified in input parameter. The method is applicable for both deterministic and stochastic analysis. The statistical attributes of model results are based on the performance function in sample of input variables. It is easier to solve complex engineering problems involving various distribution and highly non-linear engineering models.

Latin Hypercubesimulation is a recent development that gives comparable results to Monte Carlo simulation. Every input parameter range is divided into space of equivalent probability and a value is elected in random from every space inLatin Hypercubesimulation.

The simulation is much faster and more efficient compared to Monte Carlo simulation.

The computation uses specified number of iterations to determine wedge results such as maximum wedge depth, wedge weight and safety factor.

4.3 Plastic deformation

An important tunnel instability in weak and deformable rock mass is plastic deformation or squeezing. Plastic deformation was first described by Heim in 1878 during tunnelling in Alps (Shrestha, 2005). As per Shrestha (2014), there are two types of plastic deformation, namely instantaneous deformation and a long term deformation. Instantaneous deformation is the instantaneous response of rock mass under excavation of tunnel due to advancement of face (Panet, 1996). International Society for Rock Mechanics (ISRM) Commission in 1995 has defined squeezing as time dependent large deformation which occurs around tunnel, and is associated with creep caused by exceeding a limiting shear stress (Shrestha, 2005). Thus, plastic deformation can occur instantaneously or by creep- ing effect.

Instantaneous deformation

As explained in section 3.1.2, when an underground opening is excavated there is redis-

Evaluation on stability condition along the headrace tunnel of Kulekhani-III Hydroelectric Project

Master ’s thesis

Evaluation on stability condition along the headrace tunnel of Kulekhani-III Hydroelectric Project

Evaluation on stability condition along the headrace tunnel of Kulekhani-III Hydroelectric Project

Master’s thesis in Hydropower Development Supervisor: Krishna Kanta Panthi

June 2020

Norwegian University of Science and Technology Faculty of Engineering

Department of Geoscience and Petroleum

Contents

Chapter 1

Introduction

1.1 Background of study

1.2 Objective and Scope

1.3 Methodology of study

1.4 Limitations

Chapter 2

Rock and rock mass properties

2.1 Introduction

2.2 Rock mass strength and deformability

2.2.1 Intact rock strength

2.2.2 Rock mass strength

2.2.3 Rock mass deformability

2.3 Discontinuities

2.3.1 Joints in rock mass

2.3.2 Shear strength of Joints

2.3.3 Joint stiffness

2.3.4 Faults and weakness zones

2.4 Failure criteria

2.4.1 Mohr Coulomb failure criterion

2.4.2 Generalized Hoek-Brown failure criterion

2.4.3 Post failure characteristics

Chapter 3

Stresses and instabilities in tunnel

3.1 Rock stresses

3.1.1 In-situ stress in rock mass

3.1.2 Stress distribution around tunnel

3.2 Tunnel stability issues

3.2.1 Structurally controlled failure

3.2.2 Stress induced failure

Chapter 4

Instability evaluation methodologies

4.1 Block and wedge fall

4.2 Evaluation methodologies of block fall & wedge fall

4.2.1 Kinematic limit equilibrium (KLE) analysis

4.2.2 UnWedge analysis

4.3 Plastic deformation